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Steel Construction Published the List of Highly Influential Articles of 2020
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2025 Vol. 40, No. 8
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2025, 40(8): 1-11.
doi: 10.13206/j.gjgS24120202
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Abstract:
The bronze horse statue in Quanzhou Zhengchenggong Park was officially completed and opened to the public in 2004 and has been in existence for over 20 years. Due to the lack of regular inspection and maintenance of the statue's structure during long-term use， there have been multiple instances of damage and water leaks on the surface of the bronze statue， resulting in severe corrosion of the main steel structure and the secondary support system behind the copper plate. Therefore， a comprehensive inspection of the statue's interior is required. The structure was inspected using instruments such as ultrasonic thickness gauges and vernier calipers. The wall thickness of the steel pipe was reduced by about 2 mm due to corrosion， and the measured wall thickness of the ball joint was an average of 1-2 mm less than the original design wall thickness. The damaged thickness of the steel plate was 1 mm. Based on the construction drawings and actual measurement results， an overall structural evaluation model was established. The calculation results showed that the stress of the steel plate on the horse's back far exceeded the design value for steels， and the stress ratio of most members in the body and cloak was much greater than 1.0. The displacement ratio under the X-direction wind load was much greater than 1/150， indicating that the strength and stiffness of the main structure had exceeded the limit. A joint finite element analysis model was established in ABAQUS using the measured thickness， and the results showed that the stress in the body joint and the left hind leg joint had exceeded the strength design value for steels. Two semi-circular section reinforcement members were used to reinforce the original corroded steel pipe. The finite element simulation results showed that the reinforcement members and the circular steel pipe could work together to improve the compressive and bending bearing performance of the component. The steel plate was reinforced by welded plates， and finite element analysis confirmed that the stress distribution of the reinforced plate was generally consistent with that of the original steel plate； thus， the plate coukd work together with the original steel plate. Circular stiffeners were added between adjacent members and ball joints， and finite element analysis proved that the stiffeners effectively participate in load transfer， thereby reinforcing and strengthening the joints. Based on the reinforcement measures taken， establish a reinforced overall analysis model in MIDAS/Gen， and then apply the load combination conditions from the current specifications to complete the overall structural reinforcement analysis. The results indicated that the overall structural strength and stiffness indicators after implementing reinforcement measures met the requirements of the current relevant specifications. A numerical wind tunnel model was established using RWIND to obtain the wind pressure distribution of the structure under wind loads in the X and Y directions. The wind pressure was then applied to the structural analysis model for wind stability analysis. The results showed that the critical load coefficients in the X and Y directions were 10.7 and 25.9， respectively， both significantly higher than the limit of 4.2 in the Technical Specification for Space Frame Structures（JGJ 7‒2010） .
The bronze horse statue in Quanzhou Zhengchenggong Park was officially completed and opened to the public in 2004 and has been in existence for over 20 years. Due to the lack of regular inspection and maintenance of the statue's structure during long-term use， there have been multiple instances of damage and water leaks on the surface of the bronze statue， resulting in severe corrosion of the main steel structure and the secondary support system behind the copper plate. Therefore， a comprehensive inspection of the statue's interior is required. The structure was inspected using instruments such as ultrasonic thickness gauges and vernier calipers. The wall thickness of the steel pipe was reduced by about 2 mm due to corrosion， and the measured wall thickness of the ball joint was an average of 1-2 mm less than the original design wall thickness. The damaged thickness of the steel plate was 1 mm. Based on the construction drawings and actual measurement results， an overall structural evaluation model was established. The calculation results showed that the stress of the steel plate on the horse's back far exceeded the design value for steels， and the stress ratio of most members in the body and cloak was much greater than 1.0. The displacement ratio under the X-direction wind load was much greater than 1/150， indicating that the strength and stiffness of the main structure had exceeded the limit. A joint finite element analysis model was established in ABAQUS using the measured thickness， and the results showed that the stress in the body joint and the left hind leg joint had exceeded the strength design value for steels. Two semi-circular section reinforcement members were used to reinforce the original corroded steel pipe. The finite element simulation results showed that the reinforcement members and the circular steel pipe could work together to improve the compressive and bending bearing performance of the component. The steel plate was reinforced by welded plates， and finite element analysis confirmed that the stress distribution of the reinforced plate was generally consistent with that of the original steel plate； thus， the plate coukd work together with the original steel plate. Circular stiffeners were added between adjacent members and ball joints， and finite element analysis proved that the stiffeners effectively participate in load transfer， thereby reinforcing and strengthening the joints. Based on the reinforcement measures taken， establish a reinforced overall analysis model in MIDAS/Gen， and then apply the load combination conditions from the current specifications to complete the overall structural reinforcement analysis. The results indicated that the overall structural strength and stiffness indicators after implementing reinforcement measures met the requirements of the current relevant specifications. A numerical wind tunnel model was established using RWIND to obtain the wind pressure distribution of the structure under wind loads in the X and Y directions. The wind pressure was then applied to the structural analysis model for wind stability analysis. The results showed that the critical load coefficients in the X and Y directions were 10.7 and 25.9， respectively， both significantly higher than the limit of 4.2 in the Technical Specification for Space Frame Structures（JGJ 7‒2010） .
2025, 40(8): 12-19.
doi: 10.13206/j.gjgS25052602
Abstract:
The T3 terminal building of an airport consists of a central hall and two side corridors， with a total structural outline dimension of 411 m × 416 m. It is composed of a lower concrete high-rise structure and a top long-span steel structure roof. According to the architectural layout， structural joints are set up on both sides of the corridor， which coincide with the extension line of the long side of the hall， thus dividing the overall building into three structural zones. Given its dimensions （411 m in length and 200 m in width）， the middle area is a highly complex and ultra-long irregular structure. The seismic wave effect and complex and variable wind load distribution should be fully considered， and the seismic and wind-induced response analysis of the structure should be carried out with multi-point excitation.The first three modes of vibration of the structure are horizontal and vertical vibrations in two directions， exhibiting a relatively regular mode shape distribution. These modes are mainly sensitive to seismic effects in the X， Y， and Z directions. When analyzing the traveling wave effects of subsequent multi-point excitations， the main directions of the traveling waves are X and Y， and three seismic waves in the X， Y， and Z directions are simultaneously applied to the structure during each wave propagation. The peak values of the main， secondary， and vertical seismic waves are applied in a ratio of 1∶0.85∶0.65. The total proportion of structural columns with a wave effect coefficient greater than 1.0 is 37%， while that of the roof steel structures is 5%. Due to the good overall stiffness of the steel roof， all gravity loads are transmitted to the structural columns. As a result， the structural columns are more sensitive to seismic motion. The total base reaction force under multi-point excitation is significantly smaller than that under uniform excitation， with a mean ratio of 0.45 between the two. This occurs because， when the traveling wave effect is considered， the vibration phases of individual structural members are inconsistent， leading to partial mutual cancellation during the superposition of base shear forces. The influence coefficients for the structural traveling wave effect are as follows： 1.6 for corner columns， 1.5 for edge columns， and 1.4 for all other columns. The components near the support of the roof steel structure are taken as 1.4， and the rest of the components are taken as 1.2.A numerical wind tunnel analysis was conducted to obtain the wind pressure coefficients on the building surface at various wind direction angles of 0°， 45°， 90°， 135°， and 180°. The results showed a highly irregular wind pressure distribution， resulting from the roof's complex shape and the combination of wind pressure and suction. Therefore， when conducting wind-induced vibration analysis on the roof steel structure， the time-history wind pressure effects at various wind direction angles should be fully considered. Based on the results of wind tunnel numerical analysis， the time-history wind pressure at each point on the roof was extracted and converted into time-history nodal loads applied to the structural analysis model. The analysis showed that the steel roof structure exhibited the maximum vertical displacement under the action of wind pressure time-history loads.Subsequent calculations for the wind vibration coefficient were performed using vertical displacement as the control parameter. The calculated coefficients for the roof steel structure under time-history wind pressure at various wind angles ranged from 1.38 to 1.55. Based on these results， a value of 1.5 is proposed for design purposes.
The T3 terminal building of an airport consists of a central hall and two side corridors， with a total structural outline dimension of 411 m × 416 m. It is composed of a lower concrete high-rise structure and a top long-span steel structure roof. According to the architectural layout， structural joints are set up on both sides of the corridor， which coincide with the extension line of the long side of the hall， thus dividing the overall building into three structural zones. Given its dimensions （411 m in length and 200 m in width）， the middle area is a highly complex and ultra-long irregular structure. The seismic wave effect and complex and variable wind load distribution should be fully considered， and the seismic and wind-induced response analysis of the structure should be carried out with multi-point excitation.The first three modes of vibration of the structure are horizontal and vertical vibrations in two directions， exhibiting a relatively regular mode shape distribution. These modes are mainly sensitive to seismic effects in the X， Y， and Z directions. When analyzing the traveling wave effects of subsequent multi-point excitations， the main directions of the traveling waves are X and Y， and three seismic waves in the X， Y， and Z directions are simultaneously applied to the structure during each wave propagation. The peak values of the main， secondary， and vertical seismic waves are applied in a ratio of 1∶0.85∶0.65. The total proportion of structural columns with a wave effect coefficient greater than 1.0 is 37%， while that of the roof steel structures is 5%. Due to the good overall stiffness of the steel roof， all gravity loads are transmitted to the structural columns. As a result， the structural columns are more sensitive to seismic motion. The total base reaction force under multi-point excitation is significantly smaller than that under uniform excitation， with a mean ratio of 0.45 between the two. This occurs because， when the traveling wave effect is considered， the vibration phases of individual structural members are inconsistent， leading to partial mutual cancellation during the superposition of base shear forces. The influence coefficients for the structural traveling wave effect are as follows： 1.6 for corner columns， 1.5 for edge columns， and 1.4 for all other columns. The components near the support of the roof steel structure are taken as 1.4， and the rest of the components are taken as 1.2.A numerical wind tunnel analysis was conducted to obtain the wind pressure coefficients on the building surface at various wind direction angles of 0°， 45°， 90°， 135°， and 180°. The results showed a highly irregular wind pressure distribution， resulting from the roof's complex shape and the combination of wind pressure and suction. Therefore， when conducting wind-induced vibration analysis on the roof steel structure， the time-history wind pressure effects at various wind direction angles should be fully considered. Based on the results of wind tunnel numerical analysis， the time-history wind pressure at each point on the roof was extracted and converted into time-history nodal loads applied to the structural analysis model. The analysis showed that the steel roof structure exhibited the maximum vertical displacement under the action of wind pressure time-history loads.Subsequent calculations for the wind vibration coefficient were performed using vertical displacement as the control parameter. The calculated coefficients for the roof steel structure under time-history wind pressure at various wind angles ranged from 1.38 to 1.55. Based on these results， a value of 1.5 is proposed for design purposes.
2025, 40(8): 20-26.
doi: 10.13206/j.gjgS23020701
Abstract:
The film and television research base， located in the Changchun Film and Television Cultural Innovation Incubation Park， has a long span of 85 m and a short span of 61 m. The daylighting roof features a peculiar shape and is directly supported by irregularly distributed peripheral frame columns， resulting in highly complex stress conditions. Two schemes were mainly considered in the selection of daylighting roof structure， namely steel truss structure and single-layer aluminum alloy latticed shell structure. In view of the advantages of light weight， beautiful appearance， and low construction difficulty， the single-layer aluminum alloy latticed shell structure was ultimately adopted. Further structural analysis and design were carried out accordingly. Due to the complex structure of the single-layer aluminum alloy latticed daylighting roof， the MIDAS/Gen model and the RFEM model were used for structural calculations， and the accuracy of the models was verified by comparing the computation results. The calculation results showed that the first three modes of the two models were highly consistent with the first three buckling modes， and the errors in their natural vibration periods and eigenvalues were very small， indicating that the accuracy of both models met the calculation requirements. Further analysis revealed that both the natural vibration modes and the buckling modes of the daylighting roof latticed shell were characterized by global vibration and global buckling， indicating a reasonable structural arrangement. The MIDAS/Gen model was used to caary out an elastic analysis of single-layer aluminum alloy latticed shells. The results showed that the maximum stress in the members was 112 MPa， while the stress in most members remained below 80 MPa， meeting the requirements of the Code for Design of Aluminium Struectures（GB 50429‒2007）. Under the standard load combination （dead load + live load）， the maximum displacement of the structure was 119 mm， resulting in a displacement-to-span ratio of 1/512， which was less than the limit of 1/400 in the Technical Specification for Space Frame Structures（JGJ 7‒2010） . These analysis results demonstrated that the structure could satisfy both strength and stiffness requirements. The RFEM model was used to conduct a double-nonlinear stability analysis of the structure and to evaluate its stability bearing capacity. The distribution form of initial defects was based on the lowest buckling mode corresponding to each working condition， with a defect magnitude set at 1/300 of the span. After calculation， the minimum value of the load critical coefficient under each working condition was 2.5， which was greater than the limit of 2.0 in the Load Code for the Design of Building Structures（GB 50009‒2012）. The progressive collapse resistance of the project was analyzed using the component removal method. The collapse condition selected the two members that had failed first in the double nonlinear stability analysis as the initially failed members. According to the collapse analysis results， when the first two members failed， the stress level of the overall grid structure remained largely unchanged， indicating that progressive collapse did not occur. Key members were extracted for characteristic buckling analysis. The effective length coefficient， calculated as 1.47 according to Euler’s formula， was input into the overall design model to check the strength of the key members. The calculation results showed that the stress ratio of most members was less than or equal to 0.5， and the structure had sufficient safety reserves. A finite element analysis model of a typical joint was established to obtain the stress state of the joint under the action of the load design value. It was confirmed that the stress in each part met the requirements of the code and had a large safety reserve. The structural design concept of "strong joint and weak member" was satisfied.
The film and television research base， located in the Changchun Film and Television Cultural Innovation Incubation Park， has a long span of 85 m and a short span of 61 m. The daylighting roof features a peculiar shape and is directly supported by irregularly distributed peripheral frame columns， resulting in highly complex stress conditions. Two schemes were mainly considered in the selection of daylighting roof structure， namely steel truss structure and single-layer aluminum alloy latticed shell structure. In view of the advantages of light weight， beautiful appearance， and low construction difficulty， the single-layer aluminum alloy latticed shell structure was ultimately adopted. Further structural analysis and design were carried out accordingly. Due to the complex structure of the single-layer aluminum alloy latticed daylighting roof， the MIDAS/Gen model and the RFEM model were used for structural calculations， and the accuracy of the models was verified by comparing the computation results. The calculation results showed that the first three modes of the two models were highly consistent with the first three buckling modes， and the errors in their natural vibration periods and eigenvalues were very small， indicating that the accuracy of both models met the calculation requirements. Further analysis revealed that both the natural vibration modes and the buckling modes of the daylighting roof latticed shell were characterized by global vibration and global buckling， indicating a reasonable structural arrangement. The MIDAS/Gen model was used to caary out an elastic analysis of single-layer aluminum alloy latticed shells. The results showed that the maximum stress in the members was 112 MPa， while the stress in most members remained below 80 MPa， meeting the requirements of the Code for Design of Aluminium Struectures（GB 50429‒2007）. Under the standard load combination （dead load + live load）， the maximum displacement of the structure was 119 mm， resulting in a displacement-to-span ratio of 1/512， which was less than the limit of 1/400 in the Technical Specification for Space Frame Structures（JGJ 7‒2010） . These analysis results demonstrated that the structure could satisfy both strength and stiffness requirements. The RFEM model was used to conduct a double-nonlinear stability analysis of the structure and to evaluate its stability bearing capacity. The distribution form of initial defects was based on the lowest buckling mode corresponding to each working condition， with a defect magnitude set at 1/300 of the span. After calculation， the minimum value of the load critical coefficient under each working condition was 2.5， which was greater than the limit of 2.0 in the Load Code for the Design of Building Structures（GB 50009‒2012）. The progressive collapse resistance of the project was analyzed using the component removal method. The collapse condition selected the two members that had failed first in the double nonlinear stability analysis as the initially failed members. According to the collapse analysis results， when the first two members failed， the stress level of the overall grid structure remained largely unchanged， indicating that progressive collapse did not occur. Key members were extracted for characteristic buckling analysis. The effective length coefficient， calculated as 1.47 according to Euler’s formula， was input into the overall design model to check the strength of the key members. The calculation results showed that the stress ratio of most members was less than or equal to 0.5， and the structure had sufficient safety reserves. A finite element analysis model of a typical joint was established to obtain the stress state of the joint under the action of the load design value. It was confirmed that the stress in each part met the requirements of the code and had a large safety reserve. The structural design concept of "strong joint and weak member" was satisfied.
2025, 40(8): 27-34.
doi: 10.13206/j.gjgS22092601
Abstract:
The project is one of the largest large-scale stone sculptures in China， with a total height of 62 meters and a statue area of 4200 m². For the outer contour surface of the statue， it is proposed to use granite stone with a thickness of 30-50 cm， adopting a connection method that combines masonry and dry hanging. The thickness of the concrete wall connected to the stone is about 80 cm. Therefore， the weight of the stone hanging on the entire surface of the statue is very substantial. Based on the preliminary design scheme， it has been decided to adopt a special-shaped frame structure composed of steel-reinforced concrete beams and columns， a composite structural system with internal variable cross-section shear walls， and a steel structure for the arms， shoulders， and head. The review of the overrun items of the project shows that there are torsional irregularity， concave convex irregularity， sudden changes in size， and local irregularity. Considering the function and scale of the building， the seismic performance target is finally set as Grade C， with some key components raised to Grade B. The first three vibration modes of the structure were obtained from YJK and MIDAS/ Gen models. The results showed that the first three modes were basically the same： the first mode was horizontal vibration in the X-direction， the second was horizontal vibration in the Y-direction， and the third was torsional vibration in the XY plane. The vibration mode results showed that the structure exhibited reasonable layout and sufficient torsional stiffness. Through a trequent earthquake elastic calculation， the average error of the structural indices obtained from the YJK and MIDAS models was only 3.9%， indicating that the model is highly accurate. The maximum inter-story displacement angle was 1/893， which was less than the limit of 1/800 in the Technical Specification for Concrete Structures of Tall Buildings（JGJ 3-2010）. The minimum shear-weight ratio was 5.7%， which was 3.28% higher than the requirements of Code for theSeismic Design of Buildings（GB/T 50011-2010）. The minimum stiffness-to-weight ratio was 14.6， far exceeding the limit of 1.4 in JGJ 3-2010. The ratios of the anti-overturning moment to the overturning moment under frequent earthquakes in the X and Y directions were 4.17 and 3.79， respectively， and the zero-stress area was 0%. It is obvious that all performance indicators of the project met the specification requirements under frequent earthquakes. According to the performance objectives for fortification earthquakes， the key components of the project need to meet the requirements of flexural elasticity and shear elasticity under fortification earthquakes. Steel-reinforced concrete columns were adopted for shear wall corner columns and frame columns to achieve the performance objectives of shear resistance and flexural elasticity. Diagonal crossed rebars were provided in the coupling beam of the shear wall to achieve the performance goal of shear non-yielding. The stress ratio of the upper steel structure under fortification earthquakes was less than 0.8， meeting the performance objective of elasticity under fortification earthquakes. A fortification earthquake analysis was performed on the upper steel structure， the bearing reactions were extracted， and the results were used to guide the bearing design. The upper and lower chords of the upper steel structure were directly connected to the cross-section steel embedded in the steel-reinforced concrete column at the support， thus ensuring that loads from the upper steel structure were effectively transferred to the lower steel-reinforced concrete structure. SAUSAGE software was used for elastoplastic analysis under rare earthquakes in this project. The results showed that the seismic shear forces in the X and Y directions of the structure under rare earthquakes were approximately 2.3 times those under frequent earthquakes， and the overall energy dissipation capacity of the structure was better under rare earthquake elastoplastic conditions. The maximum elastoplastic inter-story drift ratios of the structure were 1/166 in the X direction and 1/204 in the Y direction， both of which were less than the limit of 1/100 in JGJ 3-2010. The bottom shear wall was slightly damaged， the coupling beam was slightly damaged， most of the upper shear wall was slightly damaged， some were moderately or severely damaged， and the coupling beam was moderately damaged. Obviously， the coupling beam was damaged before the shear wall， and the damage degree of the lower wall was lower than that of the upper wall， indicating that the layout of the shear wall was reasonable. The heavily damaged upper shear wall was reinforced with steel plates. For the space truss， the buckling analysis focuses not on its stability safety factor， but on identifying relatively weak parts and strengthening them based on the buckling mode. The buckling analysis results of the upper steel structure indicated that the first three buckling modes were local buckling， corresponding to the buckling failure of the arms and the head， respectively. The weak sections therefore required strengthening. This project employed the method of pouring a concrete layer between the steel structure and the suspended steel structure to improve the overall stiffness of the head and arm assembly. This method not only achieved the purpose of reinforcement but also provided anti-corrosion protection measures for the steel structure.To prevent damage to the connections of the external hanging stone during an earthquake and avoid the casualties caused by the falling of stones， a shaking table test was carried out on the connection performance of individual stone panels. The test results demonstrated that the connection between the stone and the main structure can withstand frequent， fortification， and rare earthquakes.
The project is one of the largest large-scale stone sculptures in China， with a total height of 62 meters and a statue area of 4200 m². For the outer contour surface of the statue， it is proposed to use granite stone with a thickness of 30-50 cm， adopting a connection method that combines masonry and dry hanging. The thickness of the concrete wall connected to the stone is about 80 cm. Therefore， the weight of the stone hanging on the entire surface of the statue is very substantial. Based on the preliminary design scheme， it has been decided to adopt a special-shaped frame structure composed of steel-reinforced concrete beams and columns， a composite structural system with internal variable cross-section shear walls， and a steel structure for the arms， shoulders， and head. The review of the overrun items of the project shows that there are torsional irregularity， concave convex irregularity， sudden changes in size， and local irregularity. Considering the function and scale of the building， the seismic performance target is finally set as Grade C， with some key components raised to Grade B. The first three vibration modes of the structure were obtained from YJK and MIDAS/ Gen models. The results showed that the first three modes were basically the same： the first mode was horizontal vibration in the X-direction， the second was horizontal vibration in the Y-direction， and the third was torsional vibration in the XY plane. The vibration mode results showed that the structure exhibited reasonable layout and sufficient torsional stiffness. Through a trequent earthquake elastic calculation， the average error of the structural indices obtained from the YJK and MIDAS models was only 3.9%， indicating that the model is highly accurate. The maximum inter-story displacement angle was 1/893， which was less than the limit of 1/800 in the Technical Specification for Concrete Structures of Tall Buildings（JGJ 3-2010）. The minimum shear-weight ratio was 5.7%， which was 3.28% higher than the requirements of Code for theSeismic Design of Buildings（GB/T 50011-2010）. The minimum stiffness-to-weight ratio was 14.6， far exceeding the limit of 1.4 in JGJ 3-2010. The ratios of the anti-overturning moment to the overturning moment under frequent earthquakes in the X and Y directions were 4.17 and 3.79， respectively， and the zero-stress area was 0%. It is obvious that all performance indicators of the project met the specification requirements under frequent earthquakes. According to the performance objectives for fortification earthquakes， the key components of the project need to meet the requirements of flexural elasticity and shear elasticity under fortification earthquakes. Steel-reinforced concrete columns were adopted for shear wall corner columns and frame columns to achieve the performance objectives of shear resistance and flexural elasticity. Diagonal crossed rebars were provided in the coupling beam of the shear wall to achieve the performance goal of shear non-yielding. The stress ratio of the upper steel structure under fortification earthquakes was less than 0.8， meeting the performance objective of elasticity under fortification earthquakes. A fortification earthquake analysis was performed on the upper steel structure， the bearing reactions were extracted， and the results were used to guide the bearing design. The upper and lower chords of the upper steel structure were directly connected to the cross-section steel embedded in the steel-reinforced concrete column at the support， thus ensuring that loads from the upper steel structure were effectively transferred to the lower steel-reinforced concrete structure. SAUSAGE software was used for elastoplastic analysis under rare earthquakes in this project. The results showed that the seismic shear forces in the X and Y directions of the structure under rare earthquakes were approximately 2.3 times those under frequent earthquakes， and the overall energy dissipation capacity of the structure was better under rare earthquake elastoplastic conditions. The maximum elastoplastic inter-story drift ratios of the structure were 1/166 in the X direction and 1/204 in the Y direction， both of which were less than the limit of 1/100 in JGJ 3-2010. The bottom shear wall was slightly damaged， the coupling beam was slightly damaged， most of the upper shear wall was slightly damaged， some were moderately or severely damaged， and the coupling beam was moderately damaged. Obviously， the coupling beam was damaged before the shear wall， and the damage degree of the lower wall was lower than that of the upper wall， indicating that the layout of the shear wall was reasonable. The heavily damaged upper shear wall was reinforced with steel plates. For the space truss， the buckling analysis focuses not on its stability safety factor， but on identifying relatively weak parts and strengthening them based on the buckling mode. The buckling analysis results of the upper steel structure indicated that the first three buckling modes were local buckling， corresponding to the buckling failure of the arms and the head， respectively. The weak sections therefore required strengthening. This project employed the method of pouring a concrete layer between the steel structure and the suspended steel structure to improve the overall stiffness of the head and arm assembly. This method not only achieved the purpose of reinforcement but also provided anti-corrosion protection measures for the steel structure.To prevent damage to the connections of the external hanging stone during an earthquake and avoid the casualties caused by the falling of stones， a shaking table test was carried out on the connection performance of individual stone panels. The test results demonstrated that the connection between the stone and the main structure can withstand frequent， fortification， and rare earthquakes.
2025, 40(8): 35-42.
doi: 10.13206/j.gjgS22121502
Abstract:
This project is a cast bronze statue with an overall shape of a high-rise cantilevered form， and its structural system is a special-shaped spatial truss structure. The statue is mainly divided into two parts： the vertical body section and the horizontal cantilevered cloak. The vertical body section has a height of 32 m and adopts a vertically stacked spatial steel truss with a story height of approximately 2 m. The cantilevered cloak has a length of 24 m and employs a horizontally arranged spatial steel truss with a width of approximately 2 m. Considering the stress characteristics and construction convenience of the members in each part， the upper and lower chords of the vertical section of the steel truss use I-shaped sections， the web members use round steel tubes， and the members of the cantilevered section of the steel truss also use round steel tubes. In view of the particularity of this structure， an overall structural analysis method combining seismic performance design and stability verification was adopted. A finite element analysis was carried out for critical joints to effectively ensure the safety of the structural design results. Firstly， the seismic performance objectives of the structure were determined according to its characteristics： the vertical members were designed to remain elastic under moderate earthquakes and non-yielding under major earthquakes， while the horizontal members were designed to remain elastic under minor earthquakes， non-yielding under moderate earthquakes， and partially yielding under major earthquakes. The results of the elastic analysis under minor earthquakes showed that the displacements of the structure met the code requirements when the effects of the cast aluminum cladding was fully considered. The equivalent elastic analysis results for moderate earthquakes showed that the stress ratios of vertical members were within 0.85， and those of most horizontal members were within 0.7， meeting the performance target for moderate earthquakes. The elastic-plastic time-history analysis results of the structure showed that the ductility coefficient of the members at the cloak reached 0.96， the maximum ductility coefficient of the body part members was 0.78， and the ductility coefficients of the bottom columns were all less than 0.2， which met the seismic performance objectives for major earthquakes. Then， the eigenvalue buckling analysis method and the dual nonlinear stability analysis method were used to carry out the stability check of the overall structure. The eigenvalue buckling analysis results showed that the first three buckling modes were concentrated in the vertical deformation buckling of the cantilevered cloak， and this buckling mode conformed to the structural characteristics of the cantilevered structure. The load cases of the dual nonlinear stability analysis were dead load + full live load （Case 1）， dead load + semi-distributed live load （Case 2）， and dead load + wind load （Case 3）， respectively. The calculation results showed that the critical load coefficients for the three cases were 2.1， 2.7， and 3.2， all of which exceeded the specification limit of 2.0. Some members in the span of the horizontal cloak and in the belly of the vertical body had entered the plastic stage， meaning that the members at the first yielding position would become the weak link in structural instability and failure. The design value of the elastic load for moderate earthquakes was adopted for the critical joints in the vertical part， while the design value of the non-yielding load for moderate earthquakes was adopted for the critical joints in the horizontal part. The finite element model of typical joints was established. Through finite element analysis， the stress state of critical joints under the design value of the fortification target load was obtained. The maximum stress of the critical joints in the vertical part was 270 MPa， and that of the critical joints in the horizontal part was 300 MPa. Both remained in the elastic stage， meeting the seismic performance objectives.
This project is a cast bronze statue with an overall shape of a high-rise cantilevered form， and its structural system is a special-shaped spatial truss structure. The statue is mainly divided into two parts： the vertical body section and the horizontal cantilevered cloak. The vertical body section has a height of 32 m and adopts a vertically stacked spatial steel truss with a story height of approximately 2 m. The cantilevered cloak has a length of 24 m and employs a horizontally arranged spatial steel truss with a width of approximately 2 m. Considering the stress characteristics and construction convenience of the members in each part， the upper and lower chords of the vertical section of the steel truss use I-shaped sections， the web members use round steel tubes， and the members of the cantilevered section of the steel truss also use round steel tubes. In view of the particularity of this structure， an overall structural analysis method combining seismic performance design and stability verification was adopted. A finite element analysis was carried out for critical joints to effectively ensure the safety of the structural design results. Firstly， the seismic performance objectives of the structure were determined according to its characteristics： the vertical members were designed to remain elastic under moderate earthquakes and non-yielding under major earthquakes， while the horizontal members were designed to remain elastic under minor earthquakes， non-yielding under moderate earthquakes， and partially yielding under major earthquakes. The results of the elastic analysis under minor earthquakes showed that the displacements of the structure met the code requirements when the effects of the cast aluminum cladding was fully considered. The equivalent elastic analysis results for moderate earthquakes showed that the stress ratios of vertical members were within 0.85， and those of most horizontal members were within 0.7， meeting the performance target for moderate earthquakes. The elastic-plastic time-history analysis results of the structure showed that the ductility coefficient of the members at the cloak reached 0.96， the maximum ductility coefficient of the body part members was 0.78， and the ductility coefficients of the bottom columns were all less than 0.2， which met the seismic performance objectives for major earthquakes. Then， the eigenvalue buckling analysis method and the dual nonlinear stability analysis method were used to carry out the stability check of the overall structure. The eigenvalue buckling analysis results showed that the first three buckling modes were concentrated in the vertical deformation buckling of the cantilevered cloak， and this buckling mode conformed to the structural characteristics of the cantilevered structure. The load cases of the dual nonlinear stability analysis were dead load + full live load （Case 1）， dead load + semi-distributed live load （Case 2）， and dead load + wind load （Case 3）， respectively. The calculation results showed that the critical load coefficients for the three cases were 2.1， 2.7， and 3.2， all of which exceeded the specification limit of 2.0. Some members in the span of the horizontal cloak and in the belly of the vertical body had entered the plastic stage， meaning that the members at the first yielding position would become the weak link in structural instability and failure. The design value of the elastic load for moderate earthquakes was adopted for the critical joints in the vertical part， while the design value of the non-yielding load for moderate earthquakes was adopted for the critical joints in the horizontal part. The finite element model of typical joints was established. Through finite element analysis， the stress state of critical joints under the design value of the fortification target load was obtained. The maximum stress of the critical joints in the vertical part was 270 MPa， and that of the critical joints in the horizontal part was 300 MPa. Both remained in the elastic stage， meeting the seismic performance objectives.
2025, 40(8): 43-50.
doi: 10.13206/j.gjgS23092301
Abstract:
This project is an airport navigation station， featuring an irregular horn-shaped building that narrows at the bottom and widens at the top. The structure is approximately 60 meters in height， with a top width of approximately 60 meters， a bottom width of approximately 27 meters， and a waist width of approximately 10 meters. Due to its considerable height and large overhang， the building has a top-heavy windward surface， resulting in a significant proportion of the wind load being applied at the top. This distribution is highly unfavorable for the wind resistance of the structure. Given the critical air traffic control function and the prominent overhang characteristics of this building， it is essential to conduct wind tunnel testing， numerical simulation of wind effects， and wind-induced vibration response analysis. Wind tunnel tests were conducted on a 1∶
This project is an airport navigation station， featuring an irregular horn-shaped building that narrows at the bottom and widens at the top. The structure is approximately 60 meters in height， with a top width of approximately 60 meters， a bottom width of approximately 27 meters， and a waist width of approximately 10 meters. Due to its considerable height and large overhang， the building has a top-heavy windward surface， resulting in a significant proportion of the wind load being applied at the top. This distribution is highly unfavorable for the wind resistance of the structure. Given the critical air traffic control function and the prominent overhang characteristics of this building， it is essential to conduct wind tunnel testing， numerical simulation of wind effects， and wind-induced vibration response analysis. Wind tunnel tests were conducted on a 1
2025, 40(8): 51-62.
doi: 10.13206/j.gjgS25051201
Abstract:
A certain hotel project has set up a high-altitude double-layer steel truss corridor between two towers， with a span of 69.5 m and a total mass of 3485 t， using Q420GJC box-section components. Due to site limitations （only a 20 m × 70 m working surface） and cost control for machinery， a steel column support platform assembly sliding technology was selected. The overall lifting was achieved after accumulating three sliding cycles. By analyzing the phased assembly sliding and overall lifting processes， and combined with construction process monitoring， the structural safety was ensured. The truss structure was divided into four sliding units， namely Unit 1， Unit 2， Unit 3， and Unit 4， and they were installed using the "cumulative sliding installation" method. A slip analysis model was established based on the slip process. Firstly， only the self-weight load was applied to extract the reaction force at each fulcrum， which was converted into frictional force and applied to the slip analysis model. The calculation results showed that the maximum vertical displacement of the truss during the sliding process was -4.8 mm， and the maximum stress ratio was 0.112. The strength and stiffness of the structure during the sliding process met the requirements. Based on the analysis results of the sliding process， the fulcrum reaction force and friction force were extracted and applied to the sliding beam analysis model. The results showed that the maximum vertical displacement occurred at the edge sliding beam， with a maximum vertical displacement of -7.3 mm. The maximum stress ratio of the supporting component was 0.789， and both the displacement and stress ratio of the sliding beam were below the limit. According to the analysis results， monitoring points were set up at locations with high stress and displacement during the sliding process of the truss and sliding beam. The monitoring results showed that the stress and displacement at each measuring point during the sliding process were lower than the calculated results.After sliding to the design position below， the synchronous lifting construction phase was carried out. This involved utilizing the surrounding tower structural columns to set up lifting brackets and install lifting equipment to serve as lifting points. Lower lifting points were established on the upper chord of the truss structure， which were then reinforced with stiffening members. After the suspension points were fine-tuned to level the structure， the formal lifting process commenced. The process was paused every 5 meters to adjust the structural alignment until it reached the design position.The lifting process analysis model and the lifting bracket analysis model were established based on the lifting process and structural layout. The results indicated that as long as the asynchronous lifting value was guaranteed not to exceed 25 mm during the lifting process， the impact of asynchronous effects on structural safety could be neglected. The maximum displacement of the structure occurred at the mid-span location， with a maximum vertical displacement of 20.8 mm. The lifting truss exhibited a maximum stress ratio of 0.269， while the stiffening members showed a maximum stress ratio of 0.723. The maximum vertical displacement and maximum stress ratio of the lifting brackets were -7.2 mm and 0.783， respectively. Both the strength and stiffness of the lifted structure and brackets met the specified requirements. According to the analysis results， monitoring points were set up at locations experiencing high stress and displacement on the truss and lifting brackets. The monitoring results showed that the stress and displacement at each measuring point during the lifting process were lower than the calculated values.
A certain hotel project has set up a high-altitude double-layer steel truss corridor between two towers， with a span of 69.5 m and a total mass of 3485 t， using Q420GJC box-section components. Due to site limitations （only a 20 m × 70 m working surface） and cost control for machinery， a steel column support platform assembly sliding technology was selected. The overall lifting was achieved after accumulating three sliding cycles. By analyzing the phased assembly sliding and overall lifting processes， and combined with construction process monitoring， the structural safety was ensured. The truss structure was divided into four sliding units， namely Unit 1， Unit 2， Unit 3， and Unit 4， and they were installed using the "cumulative sliding installation" method. A slip analysis model was established based on the slip process. Firstly， only the self-weight load was applied to extract the reaction force at each fulcrum， which was converted into frictional force and applied to the slip analysis model. The calculation results showed that the maximum vertical displacement of the truss during the sliding process was -4.8 mm， and the maximum stress ratio was 0.112. The strength and stiffness of the structure during the sliding process met the requirements. Based on the analysis results of the sliding process， the fulcrum reaction force and friction force were extracted and applied to the sliding beam analysis model. The results showed that the maximum vertical displacement occurred at the edge sliding beam， with a maximum vertical displacement of -7.3 mm. The maximum stress ratio of the supporting component was 0.789， and both the displacement and stress ratio of the sliding beam were below the limit. According to the analysis results， monitoring points were set up at locations with high stress and displacement during the sliding process of the truss and sliding beam. The monitoring results showed that the stress and displacement at each measuring point during the sliding process were lower than the calculated results.After sliding to the design position below， the synchronous lifting construction phase was carried out. This involved utilizing the surrounding tower structural columns to set up lifting brackets and install lifting equipment to serve as lifting points. Lower lifting points were established on the upper chord of the truss structure， which were then reinforced with stiffening members. After the suspension points were fine-tuned to level the structure， the formal lifting process commenced. The process was paused every 5 meters to adjust the structural alignment until it reached the design position.The lifting process analysis model and the lifting bracket analysis model were established based on the lifting process and structural layout. The results indicated that as long as the asynchronous lifting value was guaranteed not to exceed 25 mm during the lifting process， the impact of asynchronous effects on structural safety could be neglected. The maximum displacement of the structure occurred at the mid-span location， with a maximum vertical displacement of 20.8 mm. The lifting truss exhibited a maximum stress ratio of 0.269， while the stiffening members showed a maximum stress ratio of 0.723. The maximum vertical displacement and maximum stress ratio of the lifting brackets were -7.2 mm and 0.783， respectively. Both the strength and stiffness of the lifted structure and brackets met the specified requirements. According to the analysis results， monitoring points were set up at locations experiencing high stress and displacement on the truss and lifting brackets. The monitoring results showed that the stress and displacement at each measuring point during the lifting process were lower than the calculated values.
2025, 40(8): 63-66.
doi: 10.13206/j.gjgS24071525
Abstract:
This study analyzed the effective length factors of frame columns in a structure where columns and beams are connected back-to-back with bolted joints， and these connections were simplified as rotational springs between the beams and columns. Since both the columns and beams are continuous at the joint， each joint involves two unknown variables： the column rotation and the beam rotation， making such frames different from conventional frames. The results showed that the traditional effective length factor formula can be applied to calculate the effective length of columns in such frames， provided that a reduction factor is introduced to modify the combined linear stiffness of the left and right beams， and this reduced beam stiffness can be used to define the stiffness parameter.
This study analyzed the effective length factors of frame columns in a structure where columns and beams are connected back-to-back with bolted joints， and these connections were simplified as rotational springs between the beams and columns. Since both the columns and beams are continuous at the joint， each joint involves two unknown variables： the column rotation and the beam rotation， making such frames different from conventional frames. The results showed that the traditional effective length factor formula can be applied to calculate the effective length of columns in such frames， provided that a reduction factor is introduced to modify the combined linear stiffness of the left and right beams， and this reduced beam stiffness can be used to define the stiffness parameter.
